] [Thread] [Date] [Author]

[ad026(--nospam--at)hwcn.org (Paul Ransom)]

> From: "John P. Riley" <jpriley485(--nospam--at)peoplepc.com>

> Paul Ransom, P. Eng. wrote:
> I gag a little on the tie-rods approach. Check the elongation of the ties to
> develop the strength required by the frame thrust. Now, ask the metal
> building manufacturer what base deflection can be accomodated (or has been
> considered in design). Answer: "0 inches since it wasn't spec'd in the
> order."
> -------------------
> 
> Here's what I calculate for a 100' span:
> 
> Bar Stress (ksi)     Elongation
> 5                      0.207"
> 10                     0.424"
> 15                     0.621"
> 20                     0.828"
> 
> Are these elongations alarming or gagging?

The larger the ratio of clear span:eave height, the more significant
your deflection numbers become. A 100' span is a light-weight but gets
the example across. Using 60ksi rebar at full allowable is a really bad
plan.

A recent 100' span project (frame wt. 6200#) had a base thrust of 36k,
requiring at least 1.8 sq.in at 20 ksi. This requires at least 620# of
bar per frame (or 2500# required to achieve the lower stresses) plus
installation labour. How much rebar and concrete would replace this
value?

The last (20ksi/0.868") is very definitely a concern. Ideally, the
anchor bolts are in the correct location but the base has oversize
holes, so add another 1/2" +/- to each of your estimates. On a typical
PEMB (20' eave), this represents column off-plumb at h/360 before
construction tolerance.

Now extrapolate those horizontal differentials into a vertical
deflection at the ridge of the rigid frame (mock up a simple, 100' span,
straight member gable frame in MultiFrame, STAAD, or whatever you use,
and compare relative effects). That's a secondary effect and isn't
considered in design since the base deflection is restrained at 0". Some
combinations of geometry, member size and load won't be in the
significance zone.

The tie-rod approach is not wrong, you just have to make sure that you
do it right. Seems to me that the EOR is responsible to ensure this sort
of coordination between foundation design and steel design.

> Now, ask the metal building manufacturer, "Do you anticipate that it's
> possible to provide a foundation system infinitely rigid?"
> Answer:  "We don't worry about the foundations.  That's your problem."

That's true but not necessarily the limit. The manufacturer only
responds to what they have been given. Spec the building for a
differential foundation movement and have them price it accordingly. If
they can't (or don't want to) meet the spec or if you are not involved
in spec'ing the building, advise the owner of the extra cost for
foundation that would be required to become ADEQUATELY rigid (say, max.
2500# per frame, above). The owner can decide which is a more favourable
built cost based on what you can sleep with.

Heck, let them price the building both ways.

As I mentioned before, we have not (yet) started to see metal buildings
falling as a result of foundation differential movement. Some secondary
effects are inherently covered in FS or LRFD factors. But things are
changing and old rules of thumb and short cuts do not always apply,
anymore. I would be interested to hear if anyone has information about a
rigid frame collapse as the result of differential foundation
deflection.

-- 
Paul Ransom, P. Eng.
Civil/Structural/Project/International
Burlington, Ontario, Canada
<mailto:ad026(--nospam--at)hwcn.org> <http://www.hwcn.org/~ad026/civil.html>

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